Plastic-Zone-Controlled Grouting Rehabilitation of Deep Soft-Rock Chambers under Ultra-High Stress: Mechanism and Application

Abstract

To address the persistent large deformation and recurrent instability of the deep soft-rock chamber at the auxiliary shaft station of Yilan No. 3 Mine after repeated rehabilitation, this study systematically investigated the surrounding-rock instability mechanism and grouting-based synergistic rehabilitation method by integrating in-situ stress measurement, mineralogical analysis, theoretical calculation, numerical simulation, borehole imaging, grout-material testing, and field-scale industrial trials. The results show that the study area is located at a depth of approximately 700 m, where the maximum horizontal principal stress reaches 36.86 MPa, indicating an ultra-high-stress environment jointly controlled by tectonic and gravitational stresses. The surrounding rock contains relatively high clay mineral contents and exhibits pronounced water-sensitive softening and structural weakening. After excavation-induced unloading, a non-uniform plastic damage zone develops under the combined effects of principal stress redistribution and principal stress rotation. The theoretically calculated maximum depth of the dominant failure zone is approximately 2.5 m, while borehole imaging reveals a maximum fracture development depth of approximately 2.6 m in the arch-shoulder region, showing good agreement. The existing bolt-cable-steel set support system can restrict roadway contour deformation in the short term, but it cannot prevent the long-term expansion of the deep dominant failure zone. Based on the concepts of dominant failure zone coverage and structural reconstruction, a circumferential grouting rehabilitation method for forming a closed reinforcing structure is proposed. Grout-material tests indicate that a water-to-cement ratio of 0.8 and an additive dosage of 12% provide a favorable balance among injectability, stability, and long-term strength development. The field grouting parameters were determined as a grouting pressure of 8 MPa, a borehole depth of 3 m, and a borehole spacing of 1.5 m. Field-scale industrial trials show that, after grouting, the rates of floor heave and sidewall convergence in the test section were significantly reduced, and the deformation of the surrounding rock gradually stabilized. These findings provide a case-based reference for the long-term stability control and grouting rehabilitation design of deep soft-rock chambers under ultra-high in-situ stress.

1. Introduction

As coal mining continues to extend to greater depths, the service environment of underground roadways has progressively shifted from conventional shallow ground-pressure conditions to a coupled regime involving high in-situ stress, strong tectonic stress, intense excavation disturbance, and long-term rheological effects. After excavation of a deep soft-rock chamber, the original three-dimensional stress equilibrium of the surrounding rock is disturbed. Radial confinement is rapidly released, tangential stress concentration is intensified, and these processes are accompanied by principal stress reorientation, fracture propagation, and continuous development of the plastic damage zone. For soft-rock chambers with high argillaceous content, weak cementation, and pronounced water-sensitive softening, typical engineering problems include floor heave, inward sidewall squeezing, spalling at the arch shoulders, cracking of the shotcrete layer, and persistent deformation even after repeated rehabilitation. Previous studies have shown that large deformation in deep soft-rock roadways is not caused solely by insufficient rock strength, but results from the combined effects of high in-situ stress, weak lithology, well-developed structural discontinuities, degradation of support effectiveness, and time-dependent deformation. − For coal and rock masses containing structural surfaces, strain-energy accumulation and sudden energy release can further accelerate the transition from local damage to dynamic instability. Such progressive damage and weakening processes have also been characterized by integrated microseismic-electromagnetic monitoring and pore-structure/fractal analyses of coal. ,

Extensive research has been conducted on the instability mechanisms of deep soft-rock roadways, covering stress redistribution, floor-heave mechanisms, creep-induced failure, rock-support interaction, and asymmetric deformation. Li et al. clarified the deformation and failure mechanisms of deep soft-rock roadways and proposed corresponding support-control methods. Mo et al. investigated the buckling failure mechanism of roadway floors in coal mines through numerical simulation. Meng et al. examined instability mechanisms and control technologies for deep soft-rock roadways by combining numerical modeling with field engineering practice. Meanwhile, three-dimensional physical model tests, creep experiments, and time-dependent numerical analyses have shown that deformation in deep soft-rock surrounding rock exhibits a pronounced cumulative effect over time. Both the extent of the plastic zone and the failure pattern evolve continuously with the service duration of the support system. ,,− In addition, non-orthogonal failure, deviatoric stress effects, and softening damage under complex deep stress conditions can further modify fracture propagation paths, resulting in strongly heterogeneous and directional failure of the surrounding rock. − Recent CT-based studies on rock damage have further shown that microcrack initiation, branching, coalescence, and fracture connectivity can be quantitatively characterized from mesoscale and microscale perspectives, providing important methodological support for understanding the transition from localized cracking to connected fracture networks. , These studies provide an important foundation for understanding large deformation in deep soft rock. However, the continuous mechanism linking “excavation-induced unloading-principal stress rotation-directional fracture propagation-dominant failure zone formation” remains insufficiently clarified.

For the stability control of deep soft-rock chambers, technologies such as bolts, cables, steel sets, yielding support, compensation excavation, and combined support systems have been widely applied. Previous studies have shown that combined support and high-prestress anchorage systems can improve the confinement of shallow surrounding rock to a certain extent, while steel sets and yielding components can enhance the load-bearing capacity of the roadway contour and mitigate deformation in strongly squeezing ground. − However, for long-service underground spaces such as shaft stations, main haulage roadways, and large chambers, surrounding-rock failure is often not confined to the shallow loosened zone, but progressively extends into deeper weak structural zones. Conventional support systems mainly act on the roadway contour or the shallow surrounding rock, and their control effectiveness is strongly time dependent. Once the deeply fractured rock mass continues to deteriorate and develops a connected fracture network, merely increasing the strength of support components is insufficient to restore the inherent load-bearing capacity of the surrounding rock. Therefore, stability control of deep soft-rock chambers should shift from “passive component-based support” to “dominant failure zone identification and surrounding-rock structural reconstruction.”

Grouting and bolt-grouting technologies provide an important pathway for rehabilitating deeply fractured surrounding rock. Previous studies have shown that grouting can improve the integrity of fractured rock masses by filling fractures, cementing grout-rock interfaces, enhancing mechanical strength, and reducing permeability, thereby increasing the overall load-bearing capacity and deformation resistance of the surrounding rock. − Kang et al. systematically reviewed the theory, technology, and engineering application of grouting reinforcement in deep soft-rock roadways in coal mines, and emphasized that grouting reinforcement is essentially a process of rock-mass modification coupled with support synergy. Studies on grouting materials for fractured strata, damage repair of fractured rock masses, and stratified grouting reinforcement have also demonstrated that grout diffusion range, fracture-filling effectiveness, and continuity of the consolidation body are key factors governing the long-term performance of grouting. ,,− However, existing studies have mainly focused on material performance or construction parameters. Insufficient attention has been paid to the matching relationship between the grouting influence range and the dominant failure zone, as well as to the mechanical mechanism by which the surrounding rock is transformed from a discontinuous fractured medium into a continuous composite load-bearing structure after grouting.

On this basis, the present study takes the deep soft-rock chamber at the auxiliary shaft station of Yilan No. 3 Mine as the engineering background and addresses its persistent large deformation and recurrent instability after repeated rehabilitation. An integrated methodology combining in-situ stress measurement, mineralogical analysis, theoretical derivation, FLAC3D numerical simulation, borehole imaging, grout-material testing, and field-scale industrial trials was adopted to systematically investigate the instability mechanism of the surrounding rock and the grouting-based synergistic rehabilitation method for deep soft-rock chambers under ultra-high stress. The main objectives of this study are as follows: to establish a modified Mohr-Coulomb-based failure assessment criterion for soft rock that accounts for excavation-induced unloading and principal stress rotation, and thereby identify the dominant failure zone of the surrounding rock; to reveal the time-dependent degradation of the existing support system and the fracture development range through numerical simulation and borehole imaging; to propose a circumferential grouting-induced closed reinforcing structure reconstruction mechanism centered on full coverage of the dominant failure zone; and to determine the grout mix proportion and grouting parameters suitable for field application. The findings provide a theoretical basis and engineering reference for the long-term stability control of deep soft-rock chambers under ultra-high stress.

2. Engineering Background and Surrounding Rock Characteristics

2.1. Engineering Overview

The auxiliary shaft station at the Yilan No. 3 Mine is located at a depth of approximately 700 m within Paleogene soft rock strata, where the geological structure is highly complex. During the initial excavation stage, the roadway was supported using a conventional bolt-mesh-shotcrete system. However, in sections composed of heterogeneous composite rock masses, severe deformation soon developed, manifested by roof and sidewall cracking, subsidence, and pronounced floor heave. Although secondary rehabilitation measures involving bolt-mesh-shotcrete support combined with U-shaped steel sets were subsequently implemented, recurrent floor heave and renewed cracking of the roof and sidewalls were still observed in several sections. These repeated failures indicate that conventional support systems are inadequate to ensure the long-term stability of deep soft-rock chambers under such conditions, highlighting the urgent need for active reinforcement measures such as grouting-based rehabilitation. Rock sample testing from the surrounding rock of the auxiliary shaft station showed that the roof strata are composed mainly of oil shale, sandy mudstone, and carbonaceous mudstone, all of which exhibit inherently weak mechanical properties. In particular, the average uniaxial compressive strength of the oil shale is only 15.82 MPa. The superposition of weak surrounding rock properties and a high in-situ stress environment is therefore considered the principal cause of the sustained deformation and progressive failure of the chamber. The lithologic column of the roof and floor strata is shown in Figure .

1.

Stratigraphic column of the coal-bearing strata.

2.2. In-Situ Stress Measurement and Characterization

The in-situ stress field at Yilan No. 3 Mine was determined using an SYY-56 hydraulic fracturing stress measurement system. Six measurement points were selected, located respectively in the western auxiliary haulage roadway, the bypass of the shaft station, the western return airway at the floor of the middle coal seam, the western return airway at the floor of the lower coal seam, the return-air connection roadway of Panel Z101 at 20 m from the entrance to the western return airway, and the western return airway at 40 m from the junction with the return-air connection roadway of Panel 101. Based on the recorded hydraulic fracturing pressure curves, the underlying principles of the hydraulic fracturing method, and the impression orientations of the fractured boreholes, the in-situ stress results were obtained as summarized in Table .

1. In-Situ Stress Measurement Results.

Measuring station	Burial depth (m)	Vertical stress σ v  (MPa)	Maximum horizontal stress σ H  (MPa)	Minimum horizontal stress σ h  (MPa)	Lateral pressure coefficient η	Orientation of maximum horizontal stress
I	686.9	17.17	33.84	17.38	1.97	N27.7°W
II	689.9	17.25	36.86	18.89	2.14	N37.9°W
III	678.9	16.97	19.68	10.71	1.16	N23.2°W
IV	678.9	16.97	23.57	12.21	1.39	N24.5°W
V	671.0	16.78	20.60	10.07	1.23	N63.5°W
VI	679.9	17.00	20.50	10.02	1.21	N82.9°W

According to Table , the in-situ stress field at Yilan No. 3 Mine exhibits the following characteristics:

• (1)The maximum horizontal principal stresses measured at the six monitoring stations were 33.84, 36.86, 19.68, 23.57, 20.60, and 20.50 MPa, respectively. According to the commonly adopted stress classification criteria, stress levels of 0-10 MPa correspond to low-stress conditions, 10-18 MPa to moderate-stress conditions, 18-30 MPa to high-stress conditions, and values exceeding 30 MPa to ultra-high-stress conditions. On this basis, the principal stress regime at Yilan No. 3 Mine can be classified as an ultra-high-stress environment. Such a stress regime provides the fundamental mechanical condition for the subsequent asymmetric deformation and failure of the surrounding rock.
• (2)Based on the relative magnitudes of the principal stresses at each measurement point and the corresponding lateral pressure coefficients, the maximum horizontal principal stress in the study area is consistently and markedly greater than the vertical stress, indicating that the initial in-situ stress field is jointly governed by tectonic stress and gravitational stress. According to established stress-field classification schemes, the stress regime at Yilan No. 3 Mine can therefore be identified as a tectonic-gravitational type. This stress-field characteristic provides the mechanical basis for both the asymmetric deformation of the surrounding rock and the directional propagation of the butterfly-shaped plastic zone.
• (3)The orientations of the maximum horizontal principal stress at the six measurement points were N27.7°W, N37.9°W, N23.2°W, N24.5°W, N63.5°W, and N82.9°W, respectively. Overall, the maximum horizontal principal stress at Yilan No. 3 Mine is oriented predominantly in the northwestward direction. This directional feature provides an important basis for analyzing the spatial propagation and preferential development of the plastic zone.

2.3. Mineralogical Analysis of Rock Samples

The X-ray diffraction (XRD) results of four representative surrounding-rock samples are presented in Figure . The samples were collected from different positions and depths around the shaft-station area, including sandstone, oil shale, and mudstone strata. The quartz content ranges from 25.2% to 45.1%, with an average value of 36.3%. The total feldspar content, calculated as the sum of K-feldspar and albite, ranges from 36.7% to 45.7%, with an average value of 41.8%. The clay mineral content ranges from 6.0% to 29.1%, with an average value of 19.3%, indicating evident mineralogical heterogeneity among different lithologies.

2.

Mineralogical composition of four representative surrounding-rock samples determined by XRD analysis.

The oil shale and mudstone samples show relatively high clay mineral contents of 23.1% and 29.1%, respectively. In addition, the clay-mineral analysis indicates that kaolinite and illite-montmorillonite mixed-layer minerals are important components of the clay fraction. These mineralogical characteristics suggest that the surrounding rock has a clear potential for water-sensitive softening and swelling. Under high tectonic stress, such clay-rich soft rocks are prone to strength degradation, bedding-plane activation, and progressive structural weakening after water exposure, which can reduce the resistance to plastic-zone propagation and accelerate fracture coalescence. It should be noted that the XRD results were obtained from representative samples at different lithological positions, rather than from repeated parallel samples of the same stratum. Therefore, the average values reported here are used to describe the overall mineralogical tendency and sample-to-sample variability of the surrounding rock, rather than to perform a statistical comparison among individual rock layers. More systematic mineralogical testing with additional parallel samples will be required in future work to quantify the variability of clay mineral content within each lithology.

Previous studies have shown that when the contents of clay minerals, particularly kaolinite and illite-montmorillonite mixed-layer minerals, are relatively high, the surrounding rock undergoes more severe strength degradation and structural weakening upon water exposure. Under conditions of high tectonic stress, the mineral composition of the Yilan No. 3 Mine surrounding rock is expected to reduce the critical stress threshold for plastic-zone propagation, thereby accelerating the initiation and expansion of the plastic zone and further aggravating fracture coalescence and the progressive deterioration of overall rock-mass integrity.

3. Instability Mechanism and Grouting-Induced Structural Reconstruction of Deep Soft-Rock Chambers

3.1. Modified Mohr-Coulomb-Based Failure Assessment Criterion under Principal Stress Rotation

After excavation of a deep soft-rock chamber, the original three-dimensional stress equilibrium of the surrounding rock is disturbed. The formation of a free surface causes radial confinement to be released, while tangential stress becomes concentrated around the chamber boundary. For the auxiliary shaft station of Yilan No. 3 Mine, the maximum horizontal principal stress is significantly greater than the vertical stress. In addition, the surrounding rock is mainly composed of oil shale, sandy mudstone, and carbonaceous mudstone, which are characterized by low strength, intense argillization, and pronounced water-sensitive softening. Under such ultra-high-stress soft-rock conditions, the conventional Mohr–Coulomb criterion alone is insufficient to fully describe the progressive transition of the surrounding rock from local cracking to plastic damage-zone propagation after excavation-induced unloading. Therefore, it is necessary to establish a modified failure assessment criterion for soft rock that incorporates the effect of principal stress rotation and is suitable for identifying the dominant failure zone.

The shaft-station chamber is idealized as a circular roadway with an equivalent radius of a. Under the combined action of the far-field maximum horizontal principal stress σ _H and vertical stress σ _v , and considering the time-dependent support reaction p _i (t), the stress components at an arbitrary point (r,θ) in the surrounding rock can be expressed using the Kirsch solution, as given in Eq. (1):

$$\{\begin{array}{l}{\sigma }_r(r,\theta ,t)=\frac{{\sigma }_H+{\sigma }_v}{2}(1-\frac{a^2}{r^2})+\frac{{\sigma }_H-{\sigma }_v}{2}(1-\frac{4a^2}{r^2}+\frac{3a^4}{r^4})\cos 2\theta +p_i(t)\frac{a^2}{r^2}\\ {\sigma }_{\theta }(r,\theta ,t)=\frac{{\sigma }_H+{\sigma }_v}{2}(1+\frac{a^2}{r^2})-\frac{{\sigma }_H-{\sigma }_v}{2}(1+\frac{3a^4}{r^4})\cos 2\theta -p_i(t)\frac{a^2}{r^2}\\ {\tau }_{r\theta }(r,\theta ,t)=-\frac{{\sigma }_H-{\sigma }_v}{2}(1+\frac{2a^2}{r^2}-\frac{3a^4}{r^4})\sin 2\theta \end{array}$$ 1

where σ _r , σ _θ and τ _rθ denote the radial stress, tangential stress, and shear stress, respectively; σ _H is the far-field maximum horizontal principal stress; σ _v is the vertical stress; a is the equivalent roadway radius; r is the distance from an arbitrary point in the surrounding rock to the roadway center; θ is the angle between this point and the horizontal direction; and p _i (t) represents the time-dependent evolution of the support reaction. Immediately after excavation, the roadway wall can be approximated as a free surface, and thus p _i (t) = 0. After the support system becomes effective, p _i (t) > 0, representing the confinement exerted on the surrounding rock by the bolts, cables, and steel sets.

The maximum and minimum principal stresses, together with the rotation angle of the principal stress axis at any point in the surrounding rock after excavation-induced unloading, can be further derived from Eq. (1), as expressed in Eq. (2):

$$\{\begin{array}{c}{\sigma }_1^{\prime }(r,\theta ,t)=\frac{{\sigma }_r+{\sigma }_{\theta }}{2}+\sqrt{{\left(\frac{{\sigma }_{\theta }-{\sigma }_r}{2}\right)}^2+{\tau }_{r\theta }^2}\\ {\sigma }_3^{\prime }(r,\theta ,t)=\frac{{\sigma }_r+{\sigma }_{\theta }}{2}-\sqrt{{\left(\frac{{\sigma }_{\theta }-{\sigma }_r}{2}\right)}^2+{\tau }_{r\theta }^2}\\ \gamma (r,\theta ,t)=\frac{1}{2}\arctan \left(\frac{2{\tau }_{r\theta }}{{\sigma }_r-{\sigma }_{\theta }}\right)-{\gamma }_0\end{array}$$ 2

where σ ₁ and σ ₃ are the maximum and minimum principal stresses after excavation-induced unloading, respectively; γ is the rotation angle of the principal stress axis induced by excavation unloading; and γ ₀ is the initial orientation angle of the maximum principal stress before excavation. In this study, γ represents the deviation angle between the post-excavation principal stress direction and the initial maximum principal stress direction. It reflects the degree of stress-axis reorientation induced by excavation unloading and can be used to characterize the tendency of fractures to propagate from random cracking toward a preferential stress-controlled direction.

eq indicates that excavation-induced unloading changes not only the magnitudes of the principal stresses in the surrounding rock, but also the orientation of the principal stress axis through the shear stress component τ _rθ . When τ _rθ increases, or when the difference between σ _r and σ _θ changes markedly, the rotation of the principal stress axis becomes more pronounced. Under these conditions, fractures in the surrounding rock are more likely to undergo shear slip and directional propagation along the preferential stress orientation. Therefore, when evaluating the critical failure state of soft rock after excavation-induced unloading, the principal stress rotation angle γ should be further incorporated into both the stress decomposition on weak planes and the correction of strength parameters.

Let β denote the angle between the normal direction of a potential weak plane or fracture surface in the surrounding rock and the initial direction of the maximum principal stress before excavation. After excavation-induced unloading, the direction of the maximum principal stress rotates, and the effective angle between the weak-plane normal and the current maximum principal stress direction becomes β – γ. According to Mohr’s stress-circle relationship, the normal stress and shear stress acting on this weak plane can be expressed as Eq. (3):

$$\{\begin{array}{c}{\sigma }_n(r,\theta ,t)=\frac{{\sigma }_1^{\prime }+{\sigma }_3^{\prime }}{2}+\frac{{\sigma }_{\mathbf{1}}^{\prime }-{\sigma }_3^{\prime }}{2}\cos 2(\beta -\gamma )\\ \tau (r,\theta ,t)=\frac{{\sigma }_1^{\prime }-{\sigma }_3^{\prime }}{2}\sin 2(\beta -\gamma )\end{array}$$ 3

where σ _n is the normal stress acting on the weak plane, τ is the shear stress acting on the weak plane, β is the angle between the weak-plane normal and the initial direction of the maximum principal stress, and γ is the rotation angle of the principal stress axis induced by excavation-induced unloading.

Equation (3) indicates that principal stress rotation does not merely change the orientation of the stress field. More importantly, it directly alters the normal confinement and shear-driving effect acting on weak planes. When γ increases the shear stress on the weak plane while reducing the normal constraint, bedding planes, joints, and argillaceous weak surfaces become more prone to opening, shear slip, and progressive coalescence.

Considering the relatively high contents of clay minerals, such as kaolinite and illite–montmorillonite mixed-layer minerals, in the surrounding rock of the shaft station at Yilan No. 3 Mine, the rock mass is prone to strength degradation and structural weakening after water exposure. To characterize this water-sensitive softening effect, a softening damage variable D _w is introduced. Accordingly, the softened cohesion and internal friction angle can be expressed as Eq. (4):

$$\{\begin{array}{c}{c}_w=c_0(1-D_w)\\ \tan {\phi }_w=(1-D_{\phi })\tan {\phi }_0\end{array}$$ 4

where c ₀ and φ₀ denote the cohesion and internal friction angle of the surrounding rock in the unsoftened state, respectively; c _w and φ _w denote the cohesion and internal friction angle after water-sensitive softening, respectively; D _w is the cohesion-softening damage variable; and D _φ is the frictional strength degradation variable. Both variables satisfy 0 ≤ D _w , D _φ < 1.

Further considering the directional weakening effect induced by principal stress rotation, the rotation angle of the principal stress axis, γ, is incorporated into the correction of the strength parameters. The modified strength parameters can then be expressed as Eq. (5):

$$\{\begin{array}{c}{c}_{\gamma }=c_w(1-{\eta }_{c}si{n}^2\gamma )\\ \tan {\phi }_{\gamma }=\tan {\phi }_w(1-{\eta }_{\phi }si{n}^2\gamma )\end{array}$$ 5

where c _γ and φ _γ denote the equivalent cohesion and equivalent internal friction angle after accounting for both water-sensitive softening and principal stress rotation, respectively; η _c and η _φ are the directional weakening coefficients of principal stress rotation for cohesion and frictional strength, respectively.

When γ = 0, no rotation of the principal stress direction occurs, and Eq. (5) degenerates into the strength-parameter expression that considers only water-sensitive softening. As γ increases, c _γ and tan φ _γ gradually decrease, indicating that principal stress rotation weakens the ability of soft rock to resist shear slip and fracture propagation. It should be noted that D _w , D _φ, η _c , and η _φ are introduced here as phenomenological parameters for engineering failure assessment, rather than as independently calibrated constitutive constants. Specifically, D _w and D _φ describe the degradation of cohesion and frictional strength caused by water-sensitive softening, respectively. In principle, they can be determined from comparative mechanical tests on natural and water-saturated rock specimens. The directional weakening coefficients η _c and η _φ are used to characterize the additional reduction in shear resistance associated with excavation-induced principal stress rotation. Their values reflect the sensitivity of weak planes, bedding, and argillaceous discontinuities to stress-axis reorientation.

In the present study, the above parameters are used mainly to construct a modified Mohr-Coulomb-based failure assessment framework for identifying the dominant failure zone around the chamber. The purpose of this framework is not to replace a fully calibrated constitutive model, but to provide a mechanically interpretable criterion for linking excavation-induced unloading, principal stress rotation, water-sensitive softening, and plastic-zone development. Therefore, the calculated failure boundary should be understood as an engineering assessment result, which needs to be cross-validated using field observations, borehole imaging, and deformation monitoring. Based on the Mohr-Coulomb strength criterion, the critical condition for shear failure along a weak plane can be expressed as Eq. (6):

$$\tau (r,\theta ,t)=c_{\gamma }+{\sigma }_n(r,\theta ,t)\tan {\phi }_{\gamma }$$ 6

Substituting Eq. (3) into Eq. (6) yields the modified failure assessment expression for a weak plane that accounts for excavation-induced unloading, principal stress rotation, and water-sensitive softening, as expressed in Eq. (7):

$$\frac{{\sigma }_1^{\prime }-{\sigma }_3^{\prime }}{2}\sin 2(\beta -\gamma )=c_{\gamma }+[\frac{{\sigma }_1^{\prime }+{\sigma }_3^{\prime }}{2}+\frac{{\sigma }_1^{\prime }-{\sigma }_3^{\prime }}{2}\cos 2(\beta -\gamma )]\tan {\phi }_{\gamma }$$ 7

Equation (7) indicates that shear failure along weak planes in soft rock is jointly governed by three factors: first, the principal stress difference σ ₁′ – σ ₃′ induced by excavation-induced unloading; second, the combined effect of the principal stress rotation angle γ and the weak-plane orientation β; and third, the degradation of strength parameters caused by water-sensitive softening. When the principal stress difference increases, principal stress rotation enhances the shear stress acting on the weak plane, and water-sensitive softening reduces c _γ and φ _γ , the surrounding rock becomes more prone to reaching the critical failure state.

Starting from the overall yielding state of the surrounding rock, Eq. (7) can be transformed into a critical failure function in the principal-stress space, as expressed in Eq. (8):

$$F_s(r,\theta ,t)={\sigma }_1^{\prime }-{\sigma }_3^{\prime }-({\sigma }_1^{\prime }+{\sigma }_3^{\prime })\sin {\phi }_{\gamma }-2c_{\gamma }\cos {\phi }_{\gamma }$$ 8

The failure state is determined as follows: when F _s (r,θ,t) < 0, the surrounding rock remains in an elastically stable state; when F _s (r,θ,t) = 0, the surrounding rock reaches the critical failure state; and when F _s (r,θ,t) > 0, the surrounding rock enters a plastic failure state.

The proposed failure function is therefore used as a practical criterion for locating the elastoplastic boundary and defining the dominant failure zone. It does not imply that all weakening parameters have universal values for different mines or lithologies. For a specific engineering site, these parameters should ideally be calibrated using water-softening tests, triaxial unloading tests, or back-analysis based on field deformation data. When complete calibration data are unavailable, the criterion can still provide a useful comparative assessment of failure-zone development, provided that the theoretical prediction is checked against independent field evidence. In this study, the calculated dominant failure depth was subsequently compared with borehole-imaging results to evaluate the engineering rationality of the proposed assessment method.

Compared with the conventional Mohr-Coulomb criterion, the modified assessment criterion considers not only the magnitudes of the principal stresses, but also the rotation of the principal stress direction through γ. In addition, the water-sensitive softening effect is incorporated through D _w and D _φ. Therefore, this criterion can more reasonably describe the directional propagation of fractures, the progressive degradation of strength, and the asymmetric development of the plastic zone in deep soft-rock chambers after excavation. Furthermore, the stress components σ _r , σ _θ and τ _rθ calculated from Eq. (1) are substituted into Eq. (2) to obtain σ ₁′, σ ₃′, and γ. These quantities are then introduced into Eq. (8), yielding the failure function at any arbitrary point in the surrounding rock, as expressed in Eq. (9):

$$F_s(r,\theta ,t)=F_s[{\sigma }_1^{\prime }(r,\theta ,t),{\sigma }_3^{\prime }(r,\theta ,t),\gamma (r,\theta ,t),c_{\gamma },{\phi }_{\gamma }]$$ 9

When F _s (r,θ,t) = 0, the corresponding point lies on the elastoplastic boundary. Therefore, the boundary of the plastic zone in the surrounding rock after excavation-induced unloading can be expressed as Eq. (10):

$$F_s(r_p,\theta ,t)=0$$ 10

where r _p denotes the plastic-zone boundary radius after accounting for principal stress rotation and softening effects. Because the shaft station of Yilan No. 3 Mine is located in a non-uniform high horizontal stress environment, and the maximum horizontal principal stress has a distinct preferential orientation, r _p is generally not a constant. Instead, it represents a non-circular boundary that varies with θ. This boundary can be further expressed as Eq. (11):

$$r_p=r_p(\theta ,t)$$ 11

Accordingly, the thickness of the dominant failure zone in the surrounding rock can be defined as Eq. (12):

$$D_p(\theta ,t)=r_p(\theta ,t)-a$$ 12

The maximum depth of the dominant failure zone can then be expressed as Eq. (13):

$$D_{p,\max }=\underset{\theta }{\max }[r_p(\theta ,t)-a]$$ 13

Equations (10)–(13) establish a theoretical linkage from the excavation-induced unloading stress field, to the modified failure assessment criterion incorporating principal stress rotation, and further to the determination of the dominant failure zone. Based on the physical and mechanical parameters of the coal-measure rocks at Yilan No. 3 Mine, the dominant failure zone around the shaft station was recalculated using the proposed modified failure assessment criterion, which accounts for stress redistribution under excavation-induced unloading.

The straight-wall semicircular-arch roadway, with a width of 5.2 m and a height of 3.9 m, was simplified as an equivalent circular roadway, giving an equivalent radius of a = 2.35 m. The in-situ stress measurements at six stations show evident spatial variability. Station II was adopted because it exhibits the highest maximum horizontal principal stress and therefore represents the most unfavorable condition for deformation assessment and support design. This treatment provides a conservative basis for determining the dominant failure zone and grouting coverage depth. The results obtained under Station II should therefore be interpreted as a conservative assessment of the most severe condition, rather than as a uniform stress state for the entire chamber group. The most unfavorable in-situ stress condition, corresponding to measuring station II, was adopted for the calculation. The maximum horizontal principal stress and vertical stress were taken as σ _H = 36.86 MPa and σ _v = 17.25 MPa, respectively. The cohesion and internal friction angle of each lithology were then substituted into the Mohr–Coulomb-based critical failure function for solution. The results show that the thickness of the dominant failure zone varies markedly among different lithologies. The maximum depths of the dominant failure zone in sandstone, carbonaceous mudstone, sandy mudstone, oil shale, coal, and mudstone are 0.92, 0.79, 1.11, 1.29, 2.50, and 1.77 m, respectively. Among these strata, the coal seam and weak argillaceous rock layers are the most sensitive to plastic-zone propagation, with the maximum dominant failure depth reaching 2.50 m.

This agreement does not indicate that the proposed criterion is a universal predictive model for all deep soft-rock chambers. Rather, it demonstrates that the modified failure assessment framework can reasonably capture the main spatial extent of the dominant failure zone under the specific stress, lithological, and softening conditions of the Yilan No. 3 Mine. This provides a rational basis for determining the subsequent grouting coverage depth and for shifting the control strategy from shallow support reinforcement to plastic-zone-controlled structural reconstruction.

3.2. Time-Dependent Degradation of Support Effectiveness and Evolution of the Dominant Failure Zone

To further clarify the long-term control effect of the existing bolt-cable-steel set support system on the deep soft-rock chamber at the auxiliary shaft station of Yilan No. 3 Mine, and to evaluate the engineering rationality of the principal-stress-rotation-based failure assessment criterion established in Section for determining the instability range of the surrounding rock, a numerical model of the soft-rock chamber was developed using FLAC3D. The model was constructed based on a representative field cross-section. The roadway was simplified as a straight-wall semicircular-arch chamber with a width of 5.2 m and a height of 3.9 m. The support system consisted of a combined bolt-cable-steel set structure. Considering the pronounced time-dependent deformation characteristics of deep soft rock, the surrounding-rock responses after 1, 3, 6, and 12 months of support service were simulated. These simulations were conducted to evaluate the deformation-control capacity of the existing support system and the corresponding stress adjustment behavior at different service times.

The numerical model was established according to the actual geological conditions and roadway geometry of the auxiliary shaft station. The computational domain was 50 m × 60 m × 50 m in the X-, Y-, and Z-directions, respectively, where the Y-direction was taken as the roadway axis. The roadway was arranged at the central section of the model to reduce boundary effects. Considering that the dominant failure zone extended to approximately 2.5 m, the mesh was locally refined within 3 m around the roadway contour, especially in the sidewalls, arch shoulders, and floor. The mesh size gradually increased toward the far-field boundaries. The final model contained approximately 1.0 × 10⁵ zones and 2.3 × 10⁴ gridpoints. The surrounding rock was treated as a continuous equivalent rock mass. Although the in-situ rock mass contains joints, bedding planes, and local fractures, these discontinuities were equivalently reflected through calibrated mechanical parameters and time-dependent creep parameters rather than explicitly modeled as discrete interfaces. For the elastoplastic calculation stage, the Mohr–Coulomb model was adopted because it can effectively describe shear-yielding behavior of coal-measure soft rocks under engineering-scale stress redistribution. For the long-term deformation stage, the Burgers creep model was further introduced to capture the time-dependent strain accumulation of the fractured soft rock under sustained high in-situ stress. Therefore, the numerical simulation was not intended to reproduce all mesoscopic fracture processes explicitly, but to evaluate the engineering-scale evolution of deformation, stress transfer, and support effectiveness during the service period of the existing support system.

Based on the geological data, field geological investigation, and laboratory physical-mechanical test results for the coal-measure rocks at Yilan No. 3 Mine, the mechanical parameters of each stratum were equivalently calibrated by considering joint and fracture development, reduced rock-mass integrity, and scale effects in the engineering rock mass. These calibrated parameters were then used as input parameters for the FLAC3D numerical model. The Mohr-Coulomb elastoplastic constitutive model was adopted to describe the yielding and failure behavior of the surrounding rock. The physical and mechanical parameters of each rock layer are listed in Table .

2. Physical and Mechanical Parameters of Rock Strata Used in the Numerical Model.

Rock layer	Thickness (m)	Density (kg · m–3)	Cohesion (MPa)	Internal friction angle (°)	Poisson’s ratio	Elastic modulus (GPa)
Sandstone	3.9	2770	5.30	40.04	0.19	5.32
Carbonaceous mudstone	5.5	2630	1.70	35.40	0.27	27.09
Sandy mudstone	1.3	2550	0.68	33.60	0.28	26.75
Oil shale	6.8	2300	1.80	34.45	0.26	11.04
Coal	5.5	1380	0.28	29.56	0.24	2.36
Mudstone	3.4	2610	1.00	30.25	0.26	10.50

It should be noted that the parameters listed in Table are equivalent rock-mass parameters rather than direct laboratory intact-rock parameters. The laboratory test results were first used as the basic reference, and then adjusted by considering the geological survey, fracture development, bedding characteristics, reduced rock-mass integrity, and the scale effect between laboratory specimens and the engineering rock mass. In particular, the coal seam, oil shale, mudstone, and sandy mudstone were assigned relatively reduced strength and stiffness parameters to reflect their low cementation, weak structural integrity, and water-sensitive softening characteristics. This equivalent-parameter treatment is necessary because intact specimens generally overestimate the mechanical performance of fractured engineering rock masses. The relatively high elastic modulus values assigned to carbonaceous mudstone and sandy mudstone in Table were checked against the laboratory test results and geological description. These values represent the equivalent stiffness used for the numerical calculation under the present calibration scheme. These parameters should not be interpreted as universal material constants for all soft-rock roadways, but as site-specific parameters calibrated for the geological and engineering conditions of Yilan No. 3 Mine.

Recent studies on intelligent calibration of discrete-element microscopic parameters have also emphasized that numerical parameters should be carefully calibrated against macroscopic mechanical responses rather than directly transferred from laboratory tests. To account for the long-term creep behavior of soft rock under high-stress conditions, a Burgers creep model was further introduced on the basis of the elastoplastic calculation. The corresponding creep parameters are given in Table . The Burgers creep parameters were determined through a combined procedure involving reference to laboratory and engineering soft-rock creep behavior, field deformation characteristics, and repeated trial calculations. Because fractured deep soft rock contains abundant joints, bedding planes, and defects, its creep behavior at the engineering scale may differ substantially from that of intact laboratory specimens. Therefore, the creep parameters used in this study should be regarded as equivalent parameters for reproducing the time-dependent deformation trend of the chamber, rather than intrinsic constants of a single intact rock specimen.

3. Creep Parameters of the Burgers Model.

	Model parameter
	E 1  (GPa)	E 2  (GPa)	η 1  (GPa·d)	η 2  (GPa·d)
Value	4.943	64.84	351.4	16.84

In the present numerical model, a unified equivalent Burgers parameter set was adopted to characterize the overall time-dependent deformation tendency of the fractured soft-rock system. This treatment was used because the main objective of the simulation was to evaluate the degradation of the existing support effectiveness and the cumulative deformation trend of the chamber over 1, 3, 6, and 12 months, rather than to distinguish the creep response of each individual lithology. The limitation of this treatment is acknowledged.

E ₁ and η ₁ denote the elastic modulus and viscosity coefficient of the Maxwell element, while E ₂ and η ₂ denote those of the Kelvin element. By introducing these creep parameters, the numerical model can further capture the evolution of surrounding-rock deformation from rapid early adjustment to long-term accumulation under different support service durations. This provides a basis for evaluating the time-dependent control performance of the existing support system.

The initial in-situ stress state was defined using the most unfavorable measured stress condition. The maximum horizontal principal stress, minimum horizontal principal stress, and vertical stress were taken as σ _H = 36.86 MPa, σ _h = 10.02 MPa, and σ _v = 17.25 MPa, respectively. The use of measuring station II as the initial stress condition was based on the principle of unfavorable-condition design. The in-situ stress measurements at six stations show evident spatial variability, with the maximum horizontal principal stress ranging from 19.68 to 36.86 MPa. Among them, station II exhibits the highest maximum horizontal principal stress and therefore represents the most unfavorable stress environment for deformation and failure assessment. For long-service underground chambers such as shaft stations, adopting the most unfavorable measured stress condition provides a conservative basis for evaluating the control boundary of the existing support system and for determining the required grouting coverage depth. Nevertheless, the spatial variability of in-situ stress implies that the shape and extent of the plastic damage zone may vary locally within the shaft station. Therefore, the numerical results obtained under station II should be interpreted as a conservative assessment of the most severe deformation condition, rather than as a uniform stress state for the entire chamber group. Horizontal displacement constraints were applied to the left and right boundaries of the model, while a vertical displacement constraint was imposed at the bottom boundary. An equivalent in-situ stress load was applied to the upper boundary to ensure that the computational domain could reproduce the deep tectonic–gravitational stress field.

The evolution characteristics of the surrounding-rock displacement in the X- and Z-directions under different support service times are shown in Figures and , respectively. The corresponding evolution of the maximum principal stress is presented in Figure .

3.

Evolution of surrounding-rock deformation in the X-direction under different support service times.

4.

Evolution of surrounding-rock deformation in the Z-direction under different support service times.

5.

Evolution of surrounding-rock stress under different support service times.

As shown in Figure , the X-direction displacement of the surrounding rock under different support service times is mainly concentrated in the sidewalls and floor-corner regions, and it continues to expand with increasing support duration. After 1 month of support, the range of sidewall convergence is relatively limited, indicating that the bolt-cable-steel set support system can provide a certain degree of confinement to the roadway contour in the short term. After 3 months, the sidewall displacement zone gradually extends into the deeper rock mass, and the deformation mode of the sidewall evolves from shallow local compression to deeper shear slip. After 6 months, the sidewall convergence zone further expands, suggesting that the confinement capacity of the shallow support structure on the deep weak surrounding rock begins to decline. After 12 months, the X-direction displacement increases markedly, and pronounced inward squeezing of the sidewalls occurs. This indicates that the existing support system is unable to effectively suppress the long-term lateral creep deformation of the deep soft rock.

As shown in Figure , the Z-direction displacement is mainly concentrated in the floor region, and floor heave accumulates continuously with increasing support service time. After 1 month of support, floor uplift has already occurred, although its deformation range remains relatively limited. After 3 months, both the extent and magnitude of floor heave increase markedly. After 6 months, the floor displacement zone extends into the deeper surrounding rock, accompanied by the coupled development of local plastic damage and compression–shear deformation. After 12 months, floor uplift becomes further intensified and evolves into the dominant contributor to long-term large deformation of the roadway. The simulation results show that the floor heave after 1, 3, 6, and 12 months of support is approximately 170, 270, 420, and 500 mm, respectively, while the corresponding sidewall convergence is approximately 60, 90, 210, and 410 mm. These results indicate that both floor and sidewall deformation exhibit clear time-dependent accumulation, with floor heave being more pronounced. The simulated deformation pattern is consistent with the field observations before grouting rehabilitation. Field investigation showed that the shaft station experienced progressive floor heave, sidewall cracking, and arch-shoulder damage after excavation. In the severely damaged loaded-track-line section, the deformation zone was approximately 136 m long, and the floor and track uplift reached approximately 500-600 mm before rehabilitation. The numerical result of approximately 500 mm floor heave after 12 months of support therefore agrees reasonably with the order of magnitude of the observed field deformation. This comparison indicates that the model can capture the main deformation trend and the dominant deformation location of the existing support system, although it should not be regarded as a point-to-point prediction of all local deformation details.

As shown in Figure , after roadway excavation, the maximum principal stress is mainly concentrated near the sidewalls, arch shoulders, and floor. During the early stage of support, the steel sets help form a certain load-bearing arch around the roadway, thereby partially relieving the shallow stress concentration. However, as the support service time increases, the stress concentration zones in the floor and sidewalls gradually expand, and local stress is progressively transferred into the deeper surrounding rock. Under the action of high horizontal principal stress, the principal stress direction in the sidewall and arch-shoulder regions undergoes pronounced adjustment, causing fractures to propagate directionally along the preferential stress orientation. In the floor region, continuous uplift develops under the combined effects of compression-shear deformation and softening-induced creep. These results indicate that the bolt-cable-steel set support system mainly functions through “shallow confinement, contour load-bearing, and passive support.” Although it can delay the development of surrounding-rock deformation, it cannot fundamentally restore the overall load-bearing capacity of the fractured soft rock.

Therefore, the control boundary of the existing support system can be summarized as follows. Bolts and cables can enhance the confinement of the shallow surrounding rock, but they cannot effectively penetrate and control the dominant failure zone. Steel sets can bear passive loads after roadway contour convergence, but they cannot actively improve the structure of the deeply fractured surrounding rock. Under the combined effects of soft-rock creep, water-sensitive softening, and principal stress rotation, the plastic damage zone continues to expand. Accordingly, long-term stability control of the shaft station cannot rely solely on increasing the strength of support components. Instead, grouting rehabilitation must be used to cover the dominant failure zone, restore rock-mass integrity, and reconstruct a continuous load-bearing structure. In this section, the numerical simulation focuses on the pre-rehabilitation condition and the time-dependent limitation of the existing bolt-cable-steel set support system. A post-grouting numerical simulation was not included because the mechanical properties of the grouted rock mass and the grout-rock interface evolve with grout diffusion, fracture filling, cementation, and curing time, which require additional in-situ coring and laboratory testing for accurate parameter determination. Instead, the effectiveness of grouting rehabilitation is evaluated in this study through the field-scale industrial trial and deformation monitoring presented in Section . This treatment avoids introducing uncalibrated post-grouting parameters into the numerical model and ensures that the grouting effect is assessed using direct engineering evidence.

3.3. Failure Characteristics of the Surrounding Rock and Field Verification

Field investigation showed that deformation was most severe in the shaft-inset area of the auxiliary shaft station at Yilan No. 3 Mine. The main failure manifestations included inward bulging of the sidewalls, cracking of the shotcrete layer, rock spalling at the arch shoulders, and floor heave. This type of failure was not a localized instability confined to a single part of the chamber. Instead, it resulted from the combined effects of stress-path alteration, principal stress rotation, activation of weak structural planes, and progressive strength degradation of the surrounding rock under deep high horizontal stress.

To further characterize the internal fracture development of the surrounding rock, a ZKXG100 borehole imaging system was used to inspect the roof, sidewalls, and arch-shoulder regions of the roadway. The borehole imaging results are shown in Figure . On this basis, and in combination with the observed deformation patterns and the principal-stress-rotation-based failure criterion, a schematic representation of the deformation and instability mechanism of the surrounding rock at the shaft station was constructed, as shown in Figure .

6.

Borehole-imaging results showing fracture distribution and failure depth in the roadway surrounding rock.

7.

Schematic illustration of the deformation, failure characteristics, and underlying mechanism of the surrounding rock at the shaft station.

As shown in Figure , fractures are widely developed within the surrounding rock of the shaft station, and the failure distribution exhibits pronounced spatial heterogeneity. The fracture development depth is approximately 2.2 m in the roof and about 1.7 m in the sidewalls. The most intensive fracture development occurs in the arch-shoulder region, where the maximum failure depth reaches approximately 2.6 m. This result is generally consistent with the maximum dominant failure depth of approximately 2.5 m obtained from the critical failure function in Section . This agreement indicates that identifying the dominant failure zone based on the plastic-zone boundary is mechanically reasonable and practically applicable.

As shown in Figure , the deformation and failure patterns of the surrounding rock differ markedly among different parts of the chamber. The roof is mainly controlled by excavation-induced unloading and bedding-plane opening. After roadway excavation, radial confinement in the near-roof region is rapidly released, causing pre-existing bedding planes and microcracks to open first. These discontinuities subsequently undergo coupled tensile–shear propagation under high horizontal stress, eventually leading to roof subsidence, shotcrete cracking, and local separation.

The floor is constrained by the underlying rock mass, and its deformation therefore exhibits a certain lag effect. However, under lateral compression induced by high horizontal stress and the time-dependent degradation of weak argillaceous strata, compression-shear failure gradually develops, resulting in progressive floor heave. The sidewalls and arch shoulders are most strongly controlled by the maximum horizontal principal stress. After the formation of lateral free surfaces, horizontal stress release and tangential stress concentration act together, generating an intense shear-stress transformation zone near the arch shoulders. In this region, the principal stress axis rotates more significantly, and directional fracture propagation and shear slip become most pronounced. Consequently, the arch shoulders represent the most severely damaged zones observed in the field.

The field fracture distribution, numerical simulation results, and theoretical calculations consistently indicate that failure of the surrounding rock at the shaft station of Yilan No. 3 Mine is not confined to the shallow loosened zone. Instead, a dominant failure zone with a thickness of approximately 2.5–2.6 m has developed. This zone represents the region where fracture coalescence and deterioration of structural integrity are most concentrated, and it is also the critical region that must be fully covered by subsequent grouting rehabilitation. If the support or grouting range is smaller than this depth, it can provide only limited confinement to the shallow fractured rock and cannot effectively prevent the continued propagation of deep fractures. Only when the rehabilitation range fully covers the dominant failure zone and forms a continuous reinforcing structure can the control strategy shift from local support to overall load-bearing reconstruction.

3.4. Structural Reconstruction Mechanism of Circumferential Grouting

Based on the preceding theoretical analysis, numerical simulations, and field observations, the instability of the surrounding rock at the shaft station of Yilan No. 3 Mine can be essentially interpreted as a structural instability induced by the combined effects of principal stress reorientation, directional fracture propagation, water-sensitive softening, and time-dependent degradation of support effectiveness under ultra-high horizontal stress. The maximum depth of the dominant failure zone is approximately 2.5–2.6 m.

For this type of deep fractured soft rock, simply strengthening support components such as bolts, cables, and steel sets cannot fundamentally alter the inherently low strength, low stiffness, and poor integrity of the fractured rock mass. Therefore, this study proposes a structural reconstruction approach centered on circumferential grouting. In this approach, grout diffuses, fills, cements, and overlaps within the dominant failure zone, transforming the originally loosened and discontinuous fractured surrounding rock into a composite load-bearing body with structural continuity and coordinated bearing capacity. The structural reconstruction mechanism of circumferential grouting is illustrated in Figure .

8.

Schematic illustration of the reconstruction mechanism of a closed reinforcing structure formed by circumferential grouting.

As shown in Figure , the role of circumferential grouting should not be understood simply as fracture filling or local strength enhancement. Its key function is to form a closed reinforcing ring around the roadway, repair the fracture network, and reconstruct the internal load-transfer paths of the surrounding rock. Through this process, the surrounding rock is transformed from a fractured and discontinuous medium into a grout–rock cemented composite. When the grout diffusion range fully covers the dominant failure zone and effective overlap is achieved between adjacent grouting boreholes, a continuous closed reinforcing structure can be formed around the roadway. This structure improves the overall strength and shear resistance of the fractured surrounding rock. It also enhances the ability of the surrounding rock to transfer reaction forces to the support system. As a result, the interaction between the surrounding rock and the support system changes from simple contact-based load bearing to coordinated load bearing.

Based on an equivalent engineering idealization, the continuous reinforced structure formed by circumferential grouting can be regarded as a closed strengthening ring surrounding the roadway and, in mechanical terms, can be approximately modeled as a thick-walled cylinder. This idealization enables a quantitative analysis of both its load-bearing capacity and its radial confinement effect. To characterize the capacity of the closed reinforcing ring to sustain the surrounding-rock load, its inner radius is defined as a = r ₀, and its outer radius as b = r ₀ + t, where t is the effective reinforcement thickness. Assuming that the outer boundary of the reinforcing ring is subjected to an equivalent surrounding-rock pressure p ₀, while the inner boundary corresponds to the free surface of the roadway so that the internal pressure may be approximated as zero, the radial and circumferential stress distributions in the thick-walled cylinder can be expressed as Eq. (14).

$$\{\begin{array}{l}{\sigma }_r^{\prime }=A-\frac{B}{r^2}\\ {\sigma }_{\theta }^{\prime }=A+\frac{B}{r^2}\end{array}$$ 14

where σ _r is radial stress in the thick-walled cylinder and σ _θ is circumferential stress in the thick-walled cylinder.

Applying the boundary conditions σ _r (a) = 0 and σ _r (b) = –p ₀, Eq. (15) can be obtained as follows:

$$\{\begin{array}{l}A=-\frac{p_o{b}^2}{b^2-a^2}\\ B=-\frac{p_o{a}^2{b}^2}{b^2-a^2}\end{array}$$ 15

The maximum circumferential compressive stress at the inner boundary of the reinforcing ring can then be derived accordingly, as given in Eq. (16):

$${|{\sigma }_{\theta }^{\prime }|}_{\max }=\frac{2p_o{r}_o^2}{r_o^2-r_i^2}$$ 16

When this circumferential compressive stress does not exceed the allowable compressive stress of the grouted consolidation body, the reconstructed natural arch can remain in a mechanically stable state. On this basis, the upper limit of the external pressure that it can sustain can be derived, as expressed in Eq. (17).

$$p_o\le \frac{{[{\sigma }_c]}_g}{2}(1-\frac{r_i^2}{r_o^2})$$ 17

where [σ _c ] _g is allowable compressive stress of the consolidation body.

The calculations indicate that the upper limit of the external pressure that the reinforcing structure can withstand is governed primarily by two factors: the allowable compressive stress of the consolidation body and the effective reinforcement thickness. If the reinforcement thickness is insufficient to cover the dominant failure zone of approximately 2.5 m, the continuous closed structure will remain confined to the shallow loosened zone and will therefore be unable to fully mobilize its load-bearing capacity. In contrast, only when the reinforcement range fully encompasses the dominant failure zone can the structural reconstruction effect acquire genuine engineering significance.

In addition to load-bearing capacity, the long-term stability of deep soft-rock chambers also depends on whether the reinforcing structure can provide sustained radial confinement. Based on the elastic solution for a thick-walled cylinder under plane-strain conditions, the radial convergence at the inner boundary of the reinforcing ring can be expressed as Eq. (18):

$$\mathrm{\Delta }u=\frac{2(1-{\nu }_g^2)}{E_g}·\frac{p_o{r}_i{r}_o^2}{r_o^2-r_i^2}$$ 18

where E _g is elastic modulus of the consolidation body and ν _g is poisson’s ratio of the consolidation body.

By further introducing the support characteristic curve from the convergence-confinement method and defining the equivalent radial stiffness of the reinforcing structure as K _S , Eq. (19) can be obtained:

$$p_o=K_s\mathrm{\Delta }u$$ 19

By combining Eqs. (18) and (19), the expression for the equivalent radial stiffness of the reinforcing structure can be derived, as given in Eq. (20):

$$K_s=\frac{E_g(r_o^2-r_i^2)}{2(1-{\nu }_g^2)r_i{r}_o^2}$$ 20

The analysis shows that the confinement stiffness of the continuous closed reinforcing structure increases with the elastic modulus of the consolidation body and is further enhanced by greater effective reinforcement thickness. This indicates that the significance of grouting-based rehabilitation extends well beyond simple fracture filling or local strength enhancement. More fundamentally, by forming a continuous closed reinforcing structure, grouting provides a sustained confining reaction to the surrounding rock, thereby transforming the originally fractured rock mass with low stiffness into a composite structure possessing improved overall rigidity and coordinated load-bearing capacity. The formation of a closed reinforcing ring is controlled not only by the grouting layout, but also by grout performance. Grout fluidity determines the diffusion coverage, stability affects the continuity of fracture filling, and the later-age strength and stiffness determine the load-bearing capacity of the grout–rock composite. Therefore, material proportioning and grouting parameters should be jointly designed to ensure the formation of a continuous closed reinforcing structure.

3.5. Strategy for Grouting-Based Synergistic Rehabilitation

Based on the principal-stress-rotation-based failure assessment criterion under excavation-induced unloading, the numerical simulation of the time-dependent degradation of the existing support system, and the borehole imaging results, the key to surrounding-rock control at the shaft station of Yilan No. 3 Mine should not be limited to strengthening individual support components. Instead, it should shift toward systematic rehabilitation targeted at the dominant failure zone. The basic strategy is as follows: the extent of the dominant failure zone is determined through theoretical calculation and borehole imaging; the long-term control boundary of the existing support system is identified through numerical simulation; the integrity of the fractured surrounding rock is restored using circumferential grouting as the core measure; and the formation of a closed reinforcing structure is ensured through optimization of grout properties and grouting parameters. Through this integrated approach, long-term suppression of floor heave, sidewall convergence, and arch-shoulder failure can be achieved.

• (1)Full coverage of the dominant failure zone. The maximum depth of the dominant failure zone at the shaft station of Yilan No. 3 Mine is approximately 2.5–2.6 m. Therefore, the effective grouting range must exceed this failure depth. In the field scheme, the grouting borehole depth was set to 3.0 m, ensuring that the grout influence zone fully covers the dominant failure zone while providing an adequate safety margin.
• (2)Formation of a continuous closed reinforcing structure. The borehole layout, borehole spacing, grouting pressure, and grout diffusion radius should be jointly designed to ensure effective overlap between adjacent diffusion zones and to avoid the formation of isolated grout veins. Only when a continuous closed reinforcing ring is formed can the internal load-transfer paths of the surrounding rock be restored and the overall bearing capacity be improved.
• (3)Balance between injectability and consolidation enhancement. Deep soft rock commonly contains dense fracture networks and presents high injection resistance. Therefore, the grout mix should not be designed solely to achieve high strength. Instead, a balanced relationship should be achieved among flowability, stability, diffusion coverage, and long-term strength development.
• (4)Improvement of the equivalent confinement stiffness of the composite structure. The existing bolt-cable-steel set support system provides a certain degree of confinement to the shallow surrounding rock, but it cannot effectively prevent the continuous expansion of the deep plastic damage zone. Grouting rehabilitation should therefore act together with the original support system to form a composite bearing system consisting of the surrounding rock, the grouted consolidation body, and the support structure, thereby improving the long-term deformation resistance of the surrounding rock.

Accordingly, this study establishes a grouting-based synergistic rehabilitation path for deep soft-rock chambers, consisting of “dominant failure zone identification-grouting range coverage-grout property matching-closed reinforcing structure formation-coordinated bearing between surrounding rock and support.” This framework provides the theoretical basis for the subsequent grout-material tests and field-scale industrial trial.

4. Development and Performance Evaluation of Grouting Materials

For cementitious reinforcement materials, previous compression tests have shown that material composition and reinforcing phases can significantly affect energy absorption, crack evolution, acoustic-emission response, and fracture-surface characteristics. Therefore, grout-material design for fractured soft-rock rehabilitation should consider not only strength development but also injectability, stability, and the ability of the consolidated body to participate in structural load bearing. The foregoing analysis indicates that the instability of the surrounding rock in the deep soft-rock roadway at Yilan No. 3 Mine is fundamentally driven by the continuous expansion of the dominant failure zone under the coupled effects of ultra-high stress and rock softening, which ultimately leads to deterioration of rock-mass integrity and progressive deformation instability. Accordingly, grouting for stability restoration should no longer be regarded merely as a material strengthening technique, but rather as a synergistic rehabilitation measure aimed at covering the dominant failure zone, re-establishing the continuity of the surrounding rock, and enhancing the load-bearing capacity of the reinforced structure. On this basis, experimental investigations on grout mix design and optimization of grouting parameters were carried out, focusing on grout injectability, diffusion coverage, and consolidation-induced strengthening, so as to provide a sound basis for determining the field rehabilitation scheme. Ordinary Portland cement was selected as the base cementitious material because of its engineering availability, relatively low cost, and environmental compatibility. The TGK additive used in this study is a mine-developed composite cement-modifying additive. Its main function is to improve particle dispersion, regulate slurry stability, enhance injectability, and promote the mechanical performance of the consolidated grout. Although the detailed proprietary formulation of the additive is not fully disclosed, the mineralogical and chemical compositions of the cement-based grout system were characterized to support material reproducibility and performance interpretation.

4.1. Experimental Design for Optimizing the Grout Mix Proportion

After excavation of the main roadway at Yilan No. 3 Mine, several large-scale weak planes were exposed, and the surrounding rock exhibited severe fracturing and damage. In this study, ordinary Portland cement was used as the base cementitious material, and a TGK-type grouting additive independently developed by the mine was incorporated to prepare a new composite grouting material. A two-factor full-combination screening test was then conducted to evaluate the performance of the grouting material. The grouting material was prepared by mixing cement with the multifunctional cement-modifying additive at a mass ratio of 9:1. The mineralogical composition and characteristic properties of the cement and the newly developed grouting material, as determined by XRD analysis, are shown in Figure .

9.

Mineralogical and chemical compositions of Portland cement and the newly developed grouting material.

To determine the optimal water and additive dosages in the grouting material, a two-factor full-combination screening design was adopted for engineering mix-proportion optimization. Five water-to-cement ratios, namely 0.4, 0.5, 0.6, 0.8, and 1.0, and four additive dosages, namely 8%, 10%, 12%, and 14%, were combined to form 20 mix proportions. This design allowed the effects of water-to-cement ratio, additive dosage, and their combined influence on grout performance to be comparatively evaluated. The purpose of the test program was engineering mix-proportion screening rather than establishing a statistical prediction model or response-surface optimization model. The main instruments used in the experiments, together with their specific models and functions, are listed in Table .

4. Main Experimental Instruments and Their Functions.

Instrument	Model/Specification	Function
Cement mortar mixer	JJ-1	Preparation and mixing of grout slurry
Standard constant-temperature and humidity curing chamber	HWS-808	Standard curing of specimens
MTS universal testing machine	MTS/SANS CMT7000	Compressive strength testing
Rotary digital viscometer	NDJ-5S	Measurement of grout viscosity
Vicat apparatus	ISO standard	Determination of setting time

The physical properties of the grout were characterized as follows. Setting times were determined in accordance with GB/T 1346-2011. The bleeding rate was measured using a 100 mL graduated cylinder by allowing the slurry to stand undisturbed for 120 min and then calculating the volume of the separated layer. Fluidity was evaluated by the free-spread method using a truncated-cone mold, with the spread diameter taken as the indicator. For mechanical testing, 70.7 mm cubic specimens were prepared in accordance with GB/T 17671-2021 and cured under standard conditions for 7 and 28 d. Uniaxial compressive strength tests were then performed using an electro-hydraulic servo universal testing machine at a loading rate of 0.25 kN/s. The overall experimental procedure is illustrated in Figure . For each mix proportion, at least three repeated measurements or parallel specimens were prepared. The results are reported as mean values, and the standard deviation was used to evaluate data dispersion where repeated data were available. The present test program was designed for engineering mix-proportion screening rather than for establishing a statistical prediction model; therefore, the optimization was based on the comprehensive balance among injectability, stability, viscosity, and strength.

10.

Preparation and performance testing procedure of the newly developed grouting material.

4.2. Analysis of the Physical Properties of the Grouting Material

As shown in Figure , under different additive dosages, both the initial and final setting times of the newly developed grout increased with increasing water-to-cement ratio, and the two exhibited broadly similar trends. This indicates that the water-to-cement ratio is the dominant factor controlling the setting behavior of the grout. When the water-to-cement ratio increased from 0.4 to 0.6, the prolongation of setting time was relatively moderate. However, once the ratio increased to 0.8, the rate of increase became markedly more pronounced. At a water-to-cement ratio of 1.0, both the initial and final setting times rose substantially, indicating a clear retardation of the setting process under high water-content conditions.

11.

Effects of water-to-cement ratio and additive dosage on the setting time of the newly developed grout.

At a given water-to-cement ratio, both the initial and final setting times generally increased as the additive dosage rose from 8% to 14%, although the differences among dosage levels remained relatively limited. This suggests that the additive exerts a certain retarding effect on grout setting, but its influence is weaker than that of the water-to-cement ratio. As the water-to-cement ratio increases, the amount of free water in the system rises and the spacing between solid particles becomes larger, thereby weakening early hydration and interparticle structural buildup and delaying the transition of the grout from a flowable plastic state to a consolidated state. The additive, by contrast, mainly suppresses the early formation of interparticle structure through adsorption effects, which further prolongs the setting process.

As shown in Figure , under different additive dosages, the bleeding rate of the newly developed grout increased with increasing water-to-cement ratio, exhibiting a clear stage-wise pattern. When the water-to-cement ratio ranged from 0.4 to 0.6, the bleeding rate remained generally low and changed only slightly among the different groups, indicating relatively good slurry stability. Once the water-to-cement ratio increased to 0.8, the bleeding rate rose markedly. At a water-to-cement ratio of 1.0, the bleeding rate increased even more substantially, demonstrating a pronounced deterioration in grout stability under high water-content conditions.

12.

Effects of water-to-cement ratio and additive dosage on the bleeding rate of the newly developed grout.

At a given water-to-cement ratio, some variation in bleeding rate was observed among different additive dosages, but the overall trend remained consistent, indicating that the additive exerts a certain regulatory effect on bleeding behavior, although its influence is less significant than that of the water-to-cement ratio. As the water-to-cement ratio increases, the proportion of free water in the slurry rises, the interparticle supporting structure becomes progressively weaker, and the tendency for solid-liquid separation is enhanced, thereby resulting in a higher bleeding rate. Although the additive can improve the structural stability of the grout to some extent, its capacity to suppress bleeding becomes limited under conditions of high free-water content.

As shown in Figure , under different additive dosages, the fluidity of the newly developed grout increased markedly with increasing water-to-cement ratio. At relatively low water-to-cement ratios, the fluidity values of all groups remained low, and the differences among additive dosages were not pronounced. As the water-to-cement ratio increased to 0.6-0.8, the differences between dosage groups gradually became more evident, accompanied by a clear improvement in grout flowability. When the water-to-cement ratio reached 1.0, the fluidity of all groups increased substantially, indicating that the enhancement in flow performance becomes much more pronounced under high water-content conditions.

13.

Effects of water-to-cement ratio and additive dosage on the fluidity of the newly developed grout.

At a given water-to-cement ratio, the additive dosage exerted a measurable influence on grout fluidity. In particular, fluidity was generally higher at additive dosages of 10%–12%, indicating that an appropriate amount of additive is more conducive to improving grout flowability. When the dosage was further increased to 14%, however, the rate of improvement became less pronounced. This behavior can be attributed to the increase in free-water content associated with higher water-to-cement ratios, which reduces interparticle internal friction and thereby enhances slurry flow. Meanwhile, a moderate additive content can improve particle dispersion and weaken the flocculated structure of the system, whereas excessive dosage may increase the viscosity of the suspension or lead to adsorption saturation, thereby limiting further improvement in flowability.

As shown in Figure , under different additive dosages, the apparent viscosity of the newly developed grout decreased markedly with increasing water-to-cement ratio. As the water-to-cement ratio increased from 0.4 to 1.0, the grout viscosity declined substantially, indicating that the rise in free-water content progressively weakened interparticle interactions and continuously reduced the rheological resistance of the slurry.

14.

Effects of water-to-cement ratio and additive dosage on the apparent viscosity of the newly developed grout.

At a given water-to-cement ratio, the viscosity of the grout generally decreased as the additive dosage increased from 8% to 10%–12%. However, when the dosage was further increased to 14%, the viscosity showed a certain degree of rebound, exhibiting a characteristic trend of first decreasing and then increasing. This indicates that an additive content of around 12% is more favorable for reducing grout viscosity. A moderate amount of additive can improve particle dispersion and weaken the flocculated structure of the system, whereas excessive addition may enhance structural buildup within the slurry or increase the viscosity of the liquid phase, thereby causing the apparent viscosity to rise again.

4.3. Analysis of the Mechanical Properties of the Grouting Material

To clearly characterize the strength development of the newly developed grouting material under different mix proportions, uniaxial compressive strength tests were conducted on grout specimens cured for 7 and 28 d, thereby enabling a systematic evaluation of their mechanical performance. As shown in Figure , the variations in uniaxial compressive strength at 7 and 28 d exhibit essentially the same overall trend, both decreasing with increasing water-to-cement ratio. This indicates that the water-to-cement ratio is the primary controlling factor governing strength development of the grout, whereas increasing curing age mainly enhances the absolute strength level without fundamentally altering the overall distribution pattern.

15.

Effects of water-to-cement ratio and additive dosage on the uniaxial compressive strength of the newly developed grout.

At a given water-to-cement ratio, both the 7 and 28 d compressive strengths increased as the additive dosage rose from 8% to 12%, with the highest values generally occurring at around 12%. When the dosage was further increased to 14%, the strength gain became less pronounced and, in some cases, showed a slight decline, indicating the existence of an optimal additive dosage range. A moderate additive content is beneficial for improving the microstructure of the grout and promoting the filling of pores by hydration products, thereby enhancing the density and load-bearing capacity of the consolidation body. In contrast, excessive additive incorporation may weaken the sustained strength development of the material.

From the standpoint of intrinsic material performance, the grout exhibited relatively favorable strength development at a water-to-cement ratio of 0.4-0.6 with an additive dosage of approximately 12%. However, given the engineering conditions at Yilan No. 3 Mine-where the deep fractured soft rock is characterized by dense crack networks and relatively high injection resistance-the field mix design must also balance injectability, diffusion coverage, and long-term mechanical performance, rather than being determined solely on the basis of compressive strength.

Although grout specimens with lower water-to-cement ratios exhibited higher compressive strength, their relatively poor flowability limited their diffusion capacity in densely fractured soft rock. In contrast, excessively high water-to-cement ratios improved flowability but reduced slurry stability and consolidated strength. Therefore, the field mix proportion was not selected solely according to the maximum compressive strength. Considering injectability, bleeding resistance, viscosity, diffusion coverage, and long-term strength development, the water-to-cement ratio of 0.8 combined with 12% additive provided a more balanced performance for the fractured soft-rock chamber at Yilan No. 3 Mine.

4.4. Correlation Analysis of Grouting Material Properties

Based on the above experimental data, the Pearson correlation coefficient method was employed to analyze the relationships between the principal physical properties of the newly developed grouting material and its uniaxial compressive strength at different curing ages, as shown in Figure . The results reveal pronounced coupling among the various performance indices, with the water-to-cement ratio exerting the most significant influence on the overall behavior of the grout system.

16.

Pearson correlation heatmap showing the coupled relationships among grout physical properties and compressive strength.

The Pearson correlation analysis further shows that the water-to-cement ratio is significantly and positively correlated with grout fluidity, bleeding rate, and setting time, while exhibiting a significant negative correlation with the 7 and 28 d uniaxial compressive strengths. This finding confirms that the water-to-cement ratio is the dominant parameter governing the coupled evolution of the physical and mechanical properties of the newly developed grouting material. As noted in the preceding analysis of physical properties, the grout system exhibited relatively good stability at water-to-cement ratios of 0.4–0.6, whereas the mechanical performance results indicated that strength development was more favorable under comparatively lower water-to-cement conditions. This discrepancy suggests that the “optimal” water-to-cement ratio is inherently dependent on the specific performance criterion being considered. By comprehensively accounting for injectability, slurry stability, and the development of long-term mechanical properties, together with the correlation analysis results, a water-to-cement ratio of 0.8 and an additive dosage of 12% were identified as providing the most favorable overall balance of material performance for engineering application. The Pearson correlation analysis was used as an auxiliary tool to clarify the coupled relationships among the physical and mechanical properties of the grout, rather than as an independent optimization model. Its purpose was to identify the dominant factor controlling the overall material behavior and to support the comprehensive selection of the field mix proportion. The results further confirmed that the water-to-cement ratio governs the coordinated evolution of setting time, bleeding rate, fluidity, viscosity, and compressive strength.

Overall, the grout-material tests indicate that the water-to-cement ratio is the primary factor controlling the balance between injectability and mechanical performance, while the TGK additive has an optimal dosage range. A moderate additive content improves slurry dispersion and consolidated strength, whereas excessive addition produces limited additional benefit. Considering the dense fracture network and high injection resistance of the deep soft rock at Yilan No. 3 Mine, a water-to-cement ratio of 0.8 and an additive dosage of 12% were selected for field application. This mix proportion provides a favorable compromise among flowability, stability, viscosity, diffusion capacity, and long-term strength, thereby supporting the formation of a continuous grouted reinforcing structure within the dominant failure zone.

5. Field Industrial Trial of Grouting Rehabilitation

5.1. Grouting Rehabilitation Scheme

To evaluate the engineering applicability of the previously identified dominant failure zone, the proposed instability mechanism, and the optimized grouting material system, a field-scale industrial trial of grouting rehabilitation was conducted in the shaft station of Yilan No. 3 Mine. The trial focused on evaluating the effectiveness of grouting reinforcement, under the optimized material-performance conditions, in rehabilitating the damaged zone and controlling the overall deformation of the surrounding rock.

According to the analysis in Section , the maximum failure depth of the surrounding rock at the shaft station is approximately 2.5 m. Based on the material test results presented in Section , a water-to-cement ratio of 0.8 combined with an additive dosage of 12% provides a favorable balance among injectability, stability, diffusion coverage, and long-term mechanical performance. On this basis, the field grouting rehabilitation parameters were determined as follows: water-to-cement ratio 0.8, additive dosage 12%, grouting pressure 8 MPa, borehole depth 3 m, and borehole spacing 1.5 m. This parameter combination ensures that the effective grout influence zone fully covers the dominant failure zone and promotes effective overlap between adjacent grout diffusion regions, thereby facilitating the formation of a continuous closed reinforcing structure around the roadway.

The field rehabilitation trial was implemented in the section of the shaft station exhibiting the most severe deformation and damage. During construction, high-pressure grouting was used to inject the grout into the fracture network of the surrounding rock, enabling it to diffuse, fill, and cement the rock mass throughout the dominant failure zone, thereby restoring rock-mass integrity and enhancing the overall load-bearing capacity. The objective of the rehabilitation was not merely to improve local strength, but to suppress the continued development of floor heave, sidewall convergence, and arch-shoulder failure through structural reconstruction.

5.2. Rehabilitation Performance of Grouting

The field trial was conducted in the loaded-track-line section of the shaft station, where deformation was most severe. The repaired section was 136 m long, with a net cross-sectional area of 18.93 m². Before rehabilitation, the floor and track uplift reached approximately 500–600 mm, and concrete spalling and cracking were widely observed in the sidewalls and arch shoulders. After rehabilitation using the optimized grouting parameters, the structural condition of the surrounding rock was improved, and the deformation rates of floor heave and sidewall convergence were reduced.

To quantitatively evaluate the deformation-control effect, field monitoring was conducted at two representative cross sections in the repaired test section. Floor heave and sidewall convergence were selected as the principal evaluation indices, and deformation data were recorded at 5-day intervals. The monitored deformation and corresponding deformation rate are shown in Figure a and Figure b, respectively, while field photographs taken 55 and 110 d after rehabilitation are shown in Figure c and Figure d.

17.

Monitoring results of surrounding-rock deformation and field appearance after grouting rehabilitation in the shaft station.

As shown in Figure a, the surrounding-rock deformation increased rapidly during the early monitoring stage and then gradually tended to stabilize. After 120 d of monitoring, the floor heave of the 1# and 2# sections stabilized at approximately 80 and 100 mm, respectively, while the corresponding sidewall convergence stabilized at approximately 40 and 45 mm. The floor deformation was greater than the sidewall convergence, indicating that floor heave remained the dominant deformation mode after rehabilitation.

The deformation-rate curves in Figure b further show that the deformation rate decreased markedly with time. During the early stage, the maximum floor-heave rate reached approximately 4.655 mm/d, while the maximum sidewall-convergence rate was approximately 1.445 mm/d. After approximately 60–70 d, the deformation rates of all monitoring lines decreased to a low level and fluctuated only slightly, indicating that the surrounding rock gradually entered a stable deformation stage. In this study, the stabilization of deformation was evaluated based on the simultaneous reduction of the deformation rate and the gradual flattening of the cumulative deformation curves. Field photographs taken 55 and 110 d after rehabilitation, as shown in Figure c and Figure d, also indicate that the roadway surface remained intact, without obvious cracking, spalling, or renewed floor heave.

To further evaluate the deformation-reduction effect of grouting, deformation monitoring data from an adjacent ungrouted section with similar geological and stress conditions were used as an engineering control. Both the grouted test section and the ungrouted control section were monitored for 120 d, and floor heave, sidewall convergence, maximum floor-heave rate, and maximum sidewall-convergence rate were selected as comparative indicators. The comparison results are summarized in Table .

5. Comparison of Surrounding-Rock Deformation between the Grouted Test Section and the Ungrouted Control Section.

Section	Reinforcement condition	Monitoring period (d)	Final floor heave (mm)	Final sidewall convergence (mm)	Maximum floor-heave rate (mm·d–1)	Maximum sidewall convergence rate (mm·d–1)
Ungrouted control section	Existing bolt-cable-steel set support	120	542	458	26.372	10.026
Grouted test section 1#	Circumferential grouting rehabilitation	120	80	40	3.161	1.233
Grouted test section 2#	Circumferential grouting rehabilitation	120	100	45	4.655	1.445

As shown in Table , the final floor heave and sidewall convergence of the ungrouted control section reached 542 mm and 458 mm, respectively, whereas those of the grouted test sections were reduced to 80-100 mm and 40-45 mm, respectively. The maximum deformation rates also decreased substantially after grouting. The maximum floor-heave rate decreased from 26.372 mm·d^–1 in the ungrouted control section to 3.161-4.655 mm·d^–1 in the grouted sections, while the maximum sidewall-convergence rate decreased from 10.026 mm·d^–1 to 1.233-1.445 mm·d^–1. These comparisons indicate that circumferential grouting not only reduced the cumulative deformation magnitude but also accelerated the transition of the surrounding rock into a low-rate stable deformation stage.

The field trial results demonstrate that the grouting parameter combination determined on the basis of the dominant failure characteristics of the plastic zone can effectively cover the principal damaged region, improve the structural integrity of the surrounding rock, and suppress subsequent deformation development. These findings confirm the engineering applicability of the proposed instability-mechanism analysis and parameter-optimization framework in practical field conditions.

6. Discussion

• (1)Instability of deep soft-rock chambers under ultra-high in-situ stress is essentially a progressive degradation process in which the surrounding rock transforms from a continuous load-bearing medium into a fractured, low-stiffness broken rock mass under the combined effects of high tectonic stress, excavation-induced unloading, principal stress rotation, directional fracture propagation, and water-sensitive softening. At the shaft station of Yilan No. 3 Mine, the maximum horizontal principal stress is significantly greater than the vertical stress. In addition, the weak argillaceous strata are well developed and contain relatively high clay mineral contents. These conditions make the surrounding rock prone to stress concentration, activation of weak planes, and strength degradation after excavation. The maximum depth of the dominant failure zone obtained from theoretical calculation is approximately 2.5 m, while borehole imaging reveals that the fracture development range in the arch-shoulder region reaches approximately 2.6 m. The close agreement between these two results indicates that identifying the dominant failure zone of deep soft-rock chambers based on the plastic-zone boundary is mechanically reasonable and practically applicable.
• (2)From the perspective of the support mechanism, the bolt–cable–steel set system mainly provides shallow confinement and passive load bearing along the roadway contour. Although it can restrict roadway deformation in the short term, it cannot repair the deep fracture network or restore the overall stiffness and continuous load-transfer paths of the fractured surrounding rock. Numerical simulations show that, as the support service time increases, floor heave and sidewall convergence continue to accumulate, while the maximum principal stress concentration zones progressively migrate into the deeper floor, sidewall, and arch-shoulder regions. This indicates that conventional support systems have a clear time-dependent control boundary. Therefore, stability control of deep soft-rock chambers under ultra-high in-situ stress cannot rely solely on increasing the strength of support components. Instead, the control strategy should shift toward full coverage of the dominant failure zone, cementation of fracture networks, and structural reconstruction of the surrounding rock.
• (3)The primary function of circumferential grouting is not merely to fill fractures or improve local strength. Rather, grout diffusion, filling, cementation, and inter-borehole overlap within the dominant failure zone enable the formation of a continuous closed reinforcing ring. Through this process, the fractured surrounding rock is transformed from a discontinuous fractured medium into a grout-rock composite with improved overall stiffness and coordinated load-bearing capacity. The equivalent thick-walled-cylinder analysis indicates that the load-bearing capacity and radial confinement stiffness of the closed reinforcing structure are mainly governed by the strength and elastic modulus of the consolidation body, as well as by the effective reinforcement thickness. Only when the grouting range fully covers the 2.5-2.6 m dominant failure zone and forms a continuously overlapping structure can further fracture propagation be effectively restrained and the internal load-transfer paths of the surrounding rock be restored.
• (4)The grout mix proportion and grouting parameters should be designed to facilitate the formation of a closed reinforcing structure, rather than simply pursuing high strength or high flowability. Grouts with a low water-to-cement ratio generally exhibit higher strength, but their diffusion capacity is limited. In contrast, grouts with a high water-to-cement ratio show better flowability, but their stability and consolidation strength are reduced. When the water-to-cement ratio is 0.8 and the additive dosage is 12%, the grout achieves a favorable balance among injectability, stability, and long-term strength development. In the field scheme, a borehole depth of 3 m, borehole spacing of 1.5 m, and grouting pressure of 8 MPa were adopted. These parameters ensure that the grout influence range fully covers the dominant failure zone and promotes effective overlap between adjacent diffusion zones. In this way, coordinated matching among material performance, fracture diffusion, and structural reconstruction can be achieved.
• (5)The field-scale industrial trial further confirmed the engineering applicability of the proposed plastic-zone-controlled grouting rehabilitation framework. After grouting rehabilitation, the rates of floor heave and sidewall convergence in the test section of the shaft station decreased markedly, and the deformation of the surrounding rock gradually stabilized. This indicates that grouting parameters determined on the basis of dominant failure zone identification can effectively improve the structural integrity of the surrounding rock and suppress subsequent time-dependent deformation. The engineering significance of this method lies in the fact that it is not an empirical rehabilitation scheme based solely on field experience. Instead, it establishes an integrated technical pathway for deep soft-rock chamber control, linking instability mechanism identification, dominant failure zone determination, grouting-range design, closed reinforcing structure formation, and coordinated bearing between the surrounding rock and the support system. It should be emphasized that the proposed method is a case-based rehabilitation framework derived from the geological and stress conditions of Yilan No. 3 Mine. The field parameters used in this study, including a borehole depth of 3 m, borehole spacing of 1.5 m, and grouting pressure of 8 MPa, should not be regarded as universal values for all deep soft-rock chambers. What is transferable is the design logic of “dominant failure zone identification-grouting range coverage-closed structural reconstruction-deformation control.” For other mines or chamber types, the dominant failure zone and grouting parameters should be re-determined according to in-situ stress conditions, lithological assemblages, fracture development characteristics, roadway geometry, grout injectability, and field trial grouting results. In addition, this study mainly verifies the deformation-control effect of grouting rehabilitation at the engineering scale. Further research is still required to clarify the long-term degradation of the grout-rock interface, the creep behavior of the consolidation body, and the long-term bearing deterioration of the closed reinforcing ring. There are still limitations in the theoretical assessment framework proposed in this study. The water-softening variables D _w and D _φ, together with the directional weakening coefficients η _c and η _φ, were introduced to describe the coupled weakening effects of water-sensitive softening and principal stress rotation. However, due to the difficulty of obtaining complete coupled water–rock softening and unloading test data for fractured deep soft rocks, these parameters were not independently calibrated through systematic laboratory tests in the present study. Therefore, the criterion should be regarded as a modified Mohr-Coulomb-based engineering assessment method for identifying the dominant failure zone, rather than as a fully calibrated constitutive model. Future studies should conduct water-saturation tests, triaxial unloading tests under different stress paths, and back-analysis based on long-term field monitoring to quantitatively determine these weakening parameters and further improve the predictive capability of the model.

7. Conclusions

• (1)The auxiliary shaft station of Yilan No. 3 Mine is located at a burial depth of approximately 700 m, where the maximum horizontal principal stress reaches 36.86 MPa. The site is therefore characterized as an ultra-high-stress environment controlled jointly by tectonic and gravitational stresses. The surrounding rock is mainly composed of oil shale, sandy mudstone, and carbonaceous mudstone, with relatively high clay mineral contents. These lithological characteristics result in pronounced water-sensitive softening and structural weakening. The combined effects of high horizontal stress and weak lithology constitute the fundamental cause of the long-term large deformation of the shaft station.
• (2)Based on the modified failure assessment criterion incorporating principal stress rotation under excavation-induced unloading, the maximum depth of the dominant failure zone around the shaft station was determined to be approximately 2.5 m. Borehole imaging revealed that the fracture development depth in the arch-shoulder region reached approximately 2.6 m, which is generally consistent with the theoretical assessment. Although the existing bolt-cable-steel set support system can constrain roadway contour deformation in the short term, it cannot prevent the long-term expansion of the deep dominant failure zone.
• (3)A circumferential grouting-based structural reconstruction method was proposed, with full coverage of the dominant failure zone as its core principle. Through fracture filling, grout-rock cementation, and overlap between adjacent diffusion zones, circumferential grouting forms a continuous closed reinforcing ring around the roadway. This process transforms the fractured surrounding rock from a low-strength and low-stiffness medium into a composite load-bearing structure capable of coordinated mechanical response.
• (4)Grout-material tests showed that the water-to-cement ratio is the dominant parameter controlling grout injectability, stability, and strength development, while the additive has an optimal dosage range. By comprehensively considering diffusion coverage and consolidation enhancement, a water-to-cement ratio of 0.8 and an additive dosage of 12% were recommended for field application. The reasonable grouting parameter combination was determined as a grouting pressure of 8 MPa, a borehole depth of 3 m, and a borehole spacing of 1.5 m.
• (5)Field-scale industrial trials showed that, after adopting the optimized grout mix and grouting parameters, the rates of floor heave and sidewall convergence in the test section of the shaft station were significantly reduced, and the deformation of the surrounding rock gradually stabilized. These results confirm the engineering applicability of the proposed technical pathway of “dominant failure zone identification-circumferential grouting coverage-closed structural reconstruction-long-term deformation control.” However, the specific grouting parameters proposed in this study are site-dependent and should be recalibrated for other projects according to their in-situ stress state, lithology, fracture development, and trial grouting response.

The datasets generated during and/or analyzed during the current study are available from the corresponding author on reasonable request.

Peng He and Gang Li: Funding acquisition, Methodology,Validation, Formal Analysis, Investigation, Visualization, Supervision, Writing – original draft, Writing – review and editing. Zhiyong Fu: Funding acquisition, Resources, Investigation. Junpeng Zou and Tian Cai: Methodology, Conceptualization, Investigation, Formal analysis. Naizhi Gao and Shushuai Wang: Funding acquisition, Resources, Investigation, Validation, Supervision. Qinghe Yang: Resources, Investigation.

This work was supported by the National Natural Science Foundation of China (No. 52174077) and the Department of Education of Liaoning Province (Nos. LJ212510147038 and LJ212510147052).

The authors declare no competing financial interest.